STRUCTURAL SYSTEMS
RESEARCH PROJECT
Report No.
SSRP– 05/ 01
Final
SEISMIC TESTING OF PRECAST
SEGMENTAL BRIDGES
PHASE III: BRIDGE SYSTEM TEST
by
KELLY P. BURNELL
SAMI H. MEGALLY
JOSÉ I. RESTREPO
FRIEDER SEIBLE
Final Report Submitted to the California Department of
Transportation ( Caltrans) Under Contract No. 59A0337
June 2005
Department of Structural Engineering
University of California, San Diego
La Jolla, California 92093- 0085
University of California, San Diego
Department of Structural Engineering
Structural Systems Research Project
Report No. SSRP– 05/ 01
SEISMIC TESTING OF PRECAST SEGMENTAL BRIDGES
PHASE III: BRIDGE SYSTEM TEST
By
Kelly P. Burnell
Graduate Student Researcher
Sami H. Megally
Graduate Student Researcher
José I. Restrepo
Professor of Structural Engineering
Frieder Seible
Dean of the Jacobs School of Engineering
Professor of Structural Engineering
Final Report Submitted to the California Department of Transportation
( Caltrans) Under Contract No. 59A0337
Department of Structural Engineering
University of California, San Diego
La Jolla, California 92093- 0085
June 2005
ii
Technical Report Documentation Page
1. Report No.
2. Government Accession No.
3. Recipient’s Catalog No.
4. Title and Subtitle
Seismic Testing of Precast Segmental Bridges
Phase III: Bridge System Test
5. Report Date
June, 2005
6. Performing Organization Code
UCSD/ SSRP- 2005/ 01
7. Author( s)
Kelly P. Burnell, Sami H. Megally, Jose I. Restrepo, Frieder Seible
8. Performing Organization Report No.
UCSD / SSRP- 2005/ 01
9. Performing Organization Name and Address
Department of Structural Engineering
School of Engineering
10. Work Unit No. ( TRAIS)
University of California, San Diego
La Jolla, California 92093- 0085
11. Contracts or Grant No.
59A0337
12. Sponsoring Agency Name and Address
California Department of Transportation
13. Type of Report and Period Covered
Final Report - June 2003 to June 2004
Division of Engineering Services
1801 30th St., West Building MS # 9/ 5I
Sacramento, California 95807
14. Sponsoring Agency Code
15. Supplementary Notes
Prepared in cooperation with the State of California Department of Transportation.
16. Abstract
This report discusses the main findings of a test examining the seismic behavior of a precast, post- tensioned, segmental bridge
superstructure with a cast- in- place, hollow, rectangular column. The half- scale specimen modeled a prototype bridge from midspan to
midspan and down to midheight of the column. The bridge was built using the balanced cantilever method and the tendon layout of
the specimen was designed to most closely match that of the prototype segment joints nearest to the column.
The test was completed in two stages; the first stage prestressing level was designed to avoid any joint openings. This was a
validation of existing bridge design procedures in which damage is limited to the column, and the superstructure remains elastic. The
second stage involved removing some of the tendons to enable inelastic deformations of the joints in the superstructure and to impose
on the joints nearest the column a more severe loading condition. The second stage of the test was done as an introductory
examination of the performance of a bridge when damage is not limited to the column and inelastic motion is allowed in the
superstructure.
The primary objectives of the test were to investigate the response of the opening of the superstructure joints, column- superstructure
interaction, plastic hinge formation in the column, and the anticipated system failure mechanism. Extensive instrumentation was used
in the column as well as the joints of the superstructure to record the damage and performance of those bridge components.
17. Key Words
post- tensioning, precast, segmental construction, balanced
cantilever, seismic, superstructures, segment joints
18. Distribution Statement
Unlimited
19. Security Classification ( of this report)
No restrictions
20. Security Classification ( of this page)
Unclassified
21. No. of Pages
~ 125
22. Price
Form DOT F 1700.7 ( 8- 72) Reproduction of completed page authorized
iii
Disclaimer
The accuracy of the information presented in this report is the sole responsibility of the
authors. All recommendations, opinions, and conclusions presented in the report are those
of the authors, and do not necessarily express the beliefs of the California Department of
Transportation or the State of California.
iv
Acknowledgments
This research project was made possible by funding from the California Department of
Transportation under contract No. 59A0337. The input of Charly Sikorsky and others at
Caltrans was greatly appreciated.
The test was performed at the Charles Lee Powell Structural Systems Laboratory at the
University of California, San Diego. As always, the capable assistance of the lab staff was
essential to the completion of the test. In particular we acknowledge L. Berman and C.
Latham, as well as B. Parks, J. Ward, M. Dyson, A. Sherman, C. Stearns, F. Ho, and the
truly incomparable undergraduate staff.
The assistance of the following groups was greatly appreciated in the material’s
procurement and manufacture of the bridge system. F. Mazloom and Keicon Precasters for
the construction of the superstructure segments. Dywidag- Systems International, Headed
Reinforcement Corporation of California, and Sika Corporation were all generous in their
assistance and with special materials needed for the project.
v
Abstract
This report discusses the main findings of a test examining the seismic behavior of a
precast, post- tensioned, segmental bridge superstructure with a cast- in- place, hollow,
rectangular column. The half- scale specimen modeled a prototype bridge from midspan to
midspan and down to midheight of the column. The bridge was built using the balanced
cantilever method and the tendon layout of the specimen was designed to most closely
match that of the prototype segment joints nearest to the column.
The test was completed in two stages; the first stage prestressing level was designed to
avoid any joint openings. This was a validation of existing bridge design procedures in
which damage is limited to the column, and the superstructure remains elastic. The second
stage involved removing some of the tendons to enable inelastic deformations of the joints
in the superstructure and to impose on the joints nearest the column a more severe loading
condition. The second stage of the test was done as an introductory examination of the
performance of a bridge when damage is not limited to the column and inelastic motion is
allowed in the superstructure.
The primary objectives of the test were to investigate the response of the column-superstructure
interaction, opening of the superstructure joints, plastic hinge formation in
the column, and the anticipated system failure mechanism. Extensive instrumentation was
used in the column as well as the joints of the superstructure to record the damage and
performance of those bridge components.
vi
Table of Contents
Disclaimer ............................................................................................................................ iii
Acknowledgments................................................................................................................ iv
Abstract ............................................................................................................................... .. v
Table of Contents .................................................................................................................. vi
List of Figures ....................................................................................................................... ix
List of Tables ..................................................................................................................... xiii
1 Introduction................................................................................................................... 1
1.1 Precast Segmental Bridges ...................................................................... 1
1.2 Test Program............................................................................................ 3
1.3 Issues To Be Addressed in Testing .......................................................... 4
1.4 Report Outline .......................................................................................... 6
2 Test Unit Design and Details ......................................................................................... 9
2.1 Overview................................................................................................... 9
2.2 Prototype Bridge Structure ....................................................................... 9
2.3 Prototype Column................................................................................... 11
2.4 Prototype Tendon Layout ....................................................................... 12
2.5 Prototype Pier Segment.......................................................................... 14
2.6 Test Unit Design and Details .................................................................. 16
2.7 Strand Bond Area Scaling ...................................................................... 24
3 Construction and Material Properties .......................................................................... 26
3.1 Overview................................................................................................. 26
3.2 Construction of the Bridge System Test Unit .......................................... 26
3.3 Material Properties.................................................................................. 28
3.3.1 Concrete Properties......................................................................... 29
3.3.2 Steel Properties ............................................................................... 31
4 Test Setup and Instrumentation.................................................................................... 33
4.1 Overview................................................................................................. 33
4.2 Test Setup .............................................................................................. 33
vii
4.3 Loading History....................................................................................... 35
4.4 Instrumentation....................................................................................... 36
4.4.1 Mild Reinforcement Strain Gages.................................................... 36
4.4.2 Tendon Strain Gauges .................................................................... 39
4.4.3 Concrete Strain Gauges .................................................................. 39
4.4.4 Curvature Instrumentation ............................................................... 40
4.4.5 Shear Instrumentation ..................................................................... 41
4.4.6 Linear Potentiometers ..................................................................... 45
4.4.7 Inclinometers ................................................................................... 46
5 Analysis and Predictions.............................................................................................. 47
5.1 Overview................................................................................................. 47
5.2 Moment- Curvature Analysis ................................................................... 47
5.3 Column Deflection .................................................................................. 50
5.4 Prestress Losses .................................................................................... 52
5.4.1 Elastic Shortening Losses ............................................................... 53
5.4.2 Friction Losses ................................................................................ 53
5.4.3 Anchorage Seating Losses.............................................................. 53
5.4.4 Long Term Losses........................................................................... 54
5.5 Joint Opening ......................................................................................... 55
6 Test Observations......................................................................................................... 58
6.1 Overview................................................................................................. 58
6.2 Test Stage 1 ........................................................................................... 58
6.3 Test Stage 2 ........................................................................................... 62
7 Discussion of Test Results ........................................................................................... 66
7.1 Overview................................................................................................. 66
7.2 Test Stage 1 ........................................................................................... 66
7.2.1 Force- Displacement Behavior ......................................................... 66
7.2.2 Curvature......................................................................................... 67
7.2.3 Longitudinal Bar Behavior ............................................................... 68
7.2.4 Transverse Bar Behavior................................................................. 70
7.2.5 Joint Behavior.................................................................................. 72
7.3 Test Stage 2 ........................................................................................... 73
7.3.1 Force- Displacement Behavior ......................................................... 73
7.3.2 Curvature......................................................................................... 74
viii
7.3.3 Longitudinal Bar Behavior ............................................................... 75
7.3.4 Column- Superstructure Interaction.................................................. 76
7.3.5 Joint Behavior.................................................................................. 77
8 Conclusions.................................................................................................................. 80
8.1 Overview................................................................................................. 80
8.2 Joint Performance................................................................................... 80
8.3 Column Performance.............................................................................. 82
8.4 Other Remarks ....................................................................................... 83
9 Appendices................................................................................................................... 84
9.1 References ............................................................................................. 84
9.2 Construction Photos ............................................................................... 87
9.3 Additional Data ..................................................................................... 103
ix
List of Figures
Figure 2.1 Prototype bridge structure.................................................................... 10
Figure 2.2 Prototype box girder cross section....................................................... 10
Figure 2.3 Location of neutral axis in column under diagonal seismic loading...... 12
Figure 2.4 Transfer of forces to superstructure during plastic hinging of column.. 13
Figure 2.5 Shear transfer through pier segment with opened joint........................ 15
Figure 2.6 Test unit setup elevation view ( not to scale) ........................................ 17
Figure 2.7 Cross section of test unit column plastic hinge region ......................... 18
Figure 2.8 Cross section of test unit column non- plastic hinge region .................. 18
Figure 2.9 Elevation of test unit column and footing ............................................. 19
Figure 2.10 Test unit pier segment reinforcement detail ....................................... 20
Figure 2.11 Cross section of test unit segments 5 and 6 ...................................... 21
Figure 2.12 Cross section of test unit segments 1- 4 and 7- 10.............................. 21
Figure 2.13 Test unit footing reinforcement detail ................................................. 22
Figure 2.14 Test unit end block reinforcement detail ............................................ 23
Figure 2.15 Test unit tendon layout....................................................................... 24
Figure 3.1 Erection sequence of precast segments.............................................. 28
Figure 4.1 Test unit east- west elevation view ....................................................... 34
Figure 4.2 Test unit plan view ............................................................................... 34
Figure 4.3 Test unit photo ..................................................................................... 35
Figure 4.4 Loading history..................................................................................... 36
Figure 4.5 Column reinforcement strain gauge layout .......................................... 37
Figure 4.6 Column reinforcement strain gauge elevation view.............................. 38
Figure 4.7 Pier segment reinforcement strain gauge layout.................................. 38
Figure 4.8 Tendon strain gauge layout ................................................................. 39
Figure 4.9 Concrete gauge layout......................................................................... 40
Figure 4.10 Column curvature instrumentation layout........................................... 41
Figure 4.11 Equivalent diagonal deformations under flexure ................................ 42
Figure 4.12 Equivalent diagonal deformations under horizontal and vertical
expansion ...................................................................................................... 43
Figure 4.13 Diagonal deformations used to calculate shear deformation ............. 43
Figure 4.14 Shear instrumentation layout ............................................................. 45
Figure 4.15 Joint displacement potentiometer layout............................................ 46
Figure 4.16 Inclinometer device layout ................................................................. 46
x
Figure 5.1 Concrete material model...................................................................... 48
Figure 5.2 Mild steel material model ..................................................................... 48
Figure 5.3 Prestressing steel material model........................................................ 49
Figure 5.4 Column moment- curvature results ....................................................... 49
Figure 5.5 Superstructure moment- curvature results............................................ 50
Figure 5.6 Force- displacement prediction for test unit .......................................... 52
Figure 5.7 Joint opening prediction for stage 1 ..................................................... 56
Figure 5.8 Joint opening prediction for stage 2 ..................................................... 57
Figure 6.1 Marked crack pattern on north and west faces at initial displacement
level of 0.5Δy.................................................................................................. 60
Figure 6.2 Marked crack pattern on north and west faces of column at μ= 2......... 60
Figure 6.3 Marked crack pattern on north and west faces of column at μ= 4......... 61
Figure 6.4 Unmarked crack pattern on south and east faces of column at μ= 4 .... 61
Figure 6.5 Top of joint crack close up during stage 2 at μ= 4................................. 63
Figure 6.6 Spalling of column south face corner regions at μ= 5 ........................... 64
Figure 6.7 Marked crack pattern on north and west faces of column at μ= 6......... 64
Figure 6.8 Marked crack pattern on north and west faces of column at μ= 8......... 65
Figure 7.1 Stage 1 force- displacement diagram ................................................... 67
Figure 7.2 Stage 1 curvature profile...................................................................... 68
Figure 7.3 Location of gauged longitudinal bars ................................................... 69
Figure 7.4 Longitudinal bar strain profiles for stage 1 at increasing ductility levels
....................................................................................................................... 70
Figure 7.5 Location of gauged transverse bar....................................................... 71
Figure 7.6 Strain profiles of the transverse reinforcement at increasing ductility
levels.............................................................................................................. 71
Figure 7.7 Joint opening during stage 1................................................................ 72
Figure 7.8 Stage 2 force- displacement diagram ................................................... 73
Figure 7.9 Stage 2 curvature profile...................................................................... 74
Figure 7.10 Longitudinal bar strain profiles for stage 2 at increasing ductility levels
....................................................................................................................... 75
Figure 7.11 Strain of corner pier segment headed bars during stage 2 ................ 77
Figure 7.12 Joint opening during stage 2.............................................................. 78
Figure 7.13 Comparison of joint opening at μ= 4 between stages 1 and 2 ............ 79
Figure 9.1 Reinforcement and placing of the footing and pin................................ 87
Figure 9.2 Corner spirals with gauges in place ..................................................... 87
Figure 9.3 Placement of corner spirals and bottom of column reinforcement ....... 88
xi
Figure 9.4 View of wall reinforcement in column................................................... 88
Figure 9.5 Column during construction prior to pouring top portion ...................... 89
Figure 9.6 Placing forms for pouring top of column............................................... 90
Figure 9.7 Pier segment reinforcement................................................................. 91
Figure 9.8 Tendon anchor and bottom of superstructure reinforcement cage ...... 91
Figure 9.9 Steel forms used for casting superstructure segments, interchangeable
inner form....................................................................................................... 92
Figure 9.10 Tendons and reinforcement in segment............................................. 92
Figure 9.11 First completed segment ( later scrapped because the bottom blisters
are backwards) .............................................................................................. 93
Figure 9.12 Unloading and storage of segments awaiting placement................... 93
Figure 9.13 Placement of 1st pair of segments...................................................... 94
Figure 9.14 Continuing placement of 1st pair of segments .................................... 94
Figure 9.15 Pouring 3” closure gap....................................................................... 95
Figure 9.16 1st pair of segments in place and epoxied with temporary prestressing
bars................................................................................................................ 96
Figure 9.17 Tendon wedge plate and wedges ...................................................... 97
Figure 9.18 Stressing the tendon .......................................................................... 97
Figure 9.19 Cutting off tendon tails and placement of grout caps ......................... 98
Figure 9.20 2nd pair of segments installed............................................................. 98
Figure 9.21 After all segments have been installed and top tendons stressed ..... 99
Figure 9.22 Placement of end blocks and fixtures ................................................ 99
Figure 9.23 Test during placement of actuators.................................................. 100
Figure 9.24 Damaged test unit after final testing cycles...................................... 101
Figure 9.25 Spalling at top corners at conclusion of testing................................ 101
Figure 9.26 Column displacement and pin rotation close- up .............................. 102
Figure 9.27 Concrete top surface strains for stage 1 .......................................... 103
Figure 9.28 Concrete bottom surface strains for stage 1 .................................... 104
Figure 9.29 Span deflections .............................................................................. 105
Figure 9.30 Column bottom rotations.................................................................. 105
Figure 9.31 Transverse bar strains at location 5A for stage 1............................. 106
Figure 9.32 Transverse bar strains at location 5A for stage 2............................. 107
Figure 9.33 Transverse bar strains at location 5F for stage 1............................. 108
Figure 9.34 Transverse bar strains at location 5F for stage 2............................. 109
Figure 9.35 Shear deflection during stage 1 ....................................................... 110
Figure 9.36 Tendon gauge microstrains for stage 1............................................ 111
xii
Figure 9.37 Tendon gauge microstrains for stage 2............................................ 112
xiii
List of Tables
Table 2- 1 Minimum number of tendons required at each joint .............................. 14
Table 3- 1 Test unit concrete compressive cylinder strengths ( ksi)....................... 30
Table 3- 2 Test unit segment compressive day of test strengths ( ksi) ................... 30
Table 3- 3 Test unit concrete compressive cylinder strengths ( MPa).................... 31
Table 3- 4 Test unit segment compressive day of test strengths ( MPa) ................ 31
Table 3- 5 Test unit mild steel reinforcement properties ( ksi)................................ 32
Table 3- 6 Test unit mild steel reinforcement properties ( MPa)............................. 32
Table 5- 1 Specimen theoretical prestress losses in ksi ( MPa).............................. 55
1
1 Introduction
1.1 Precast Segmental Bridges
Since beginning in Western Europe over fifty years ago precast segmental construction has
spread throughout the world, and in the 1970s the United States. The benefits of segmental
construction are most apparent in areas where conventional falsework is prohibitive,
shorter construction times are needed, or higher quality construction is desired. One area
in which segmental construction is still used sparingly is areas prone to seismic activity.
The process of building precast segmental bridges begins with the production of
components, or segments, of the bridge in a precasting plant or factory. The superstructure
segments are then shipped to the site of the bridge and lifted into place. Using high
strength steel tendons, the segments are post- tensioned together. Two methods of
assembling the superstucture segments dominate segmental construction. For shorter span
lengths ( 75’- 200’ ( 23- 70m)) span- by- span construction is used, and in longer spans ( 150’-
500’ ( 46- 152m)) the balanced cantilever method. The span- by- span method entails
connecting an entire span of the bridge and then lifting the span into place or assembling it
in place atop a temporary steel support. A bridge constructed by the balanced cantilever
method will be erecting by placing one segment at a time on either side of the bridge
column, thus gradually reaching out across the span distance from the column.
2
Several advantages are inherent to both building methods. Because of the modular nature
of these bridges there is no need for the extensive formwork used in conventional cast- in-place
( CIP) concrete. This is an enormous advantage in areas such as high valleys and
rivers where formwork is difficult and expensive. Also a minimal footprint is left on the
environment below the bridge whether it is an environmentally sensitive habitat or a busy
surface street. The savings accumulated by not shutting down streets located below the
bridge can be significant, especially in urban areas with high traffic flow.
Other advantages to segmental construction can be achieved as a result of precasting the
segments. Because the pieces are cast at a central fabrication facility where mix designs
and curing conditions are more easily controlled, higher quality control of segments is
achieved. Several construction tasks can also be achieved simultaneously allowing for a
reduction in the bridge’s construction schedule. For example superstructure segments can
be cast at the precast plant while preparation for and construction of piers and abutments
occurs at the construction site.
Despite the many advantages to precast segmental construction its use in earthquake- prone
areas of the United States has been hampered due to a lack of research on segmental
bridges’ performance under seismic loads. One concern is for the joints where the
segments connect. Because of a lack of experimental research in this area, very
conservative design guidelines not based on actual research currently govern the joint
region. Some previous work at the University of Texas at Austin dealt with segmental
bridges but did not address seismic issues [ 1,2] The current American Associtation of State
Highway and Transportation Officials ( AASHTO) guidelines [ 3,4] require segmental
3
bridge adhesive ( high- strength epoxy) be used on the joints as well as limiting the number
of external or unbonded tendons to 50 percent of the total prestressing in the
superstructure. The California Department of Transportation ( Caltrans) requires that mild
steel cross the joint requiring a CIP closure joint and negating many of the benefits of the
segmental construction process [ 5].
1.2 Test Program
A test program examining issues involving the use of precast segmental bridge structures
in seismic areas began in 2000 at the University of California, San Diego ( UCSD). This
program was a cooperative effort involving Caltrans, the American Segmental Bridge
Institute ( ASBI), and UCSD. The first two phases of the program looked at specific
segment to segment connections [ 6]. The first phase examined joints where high moments
and low shears are present such as the midspan of a bridge. Joints with high moments as
well as high shears like those nearest to a column were examined in phase two. The early
phases tested a series of joints with bonded tendons, unbonded tendons, and a mix of
bonded and unbonded. One notable finding was the ability of the joints to open
significantly and close with very little residual displacement or damage to the structure
when undergoing single direction loading. Further testing involving bi- directional loading
is also needed.
The third and final large- scale testing phase, and the subject of this report, was a segmental
bridge system test modeling a complete superstructure with pier segment and column.
This was done to examine issues relating to the interaction of the column and
superstructure as well as corroborating the findings of the earlier phases of the test
4
program. The prototype bridge modeled had span lengths of 100’ ( 30.5m) and was
assumed to be constructed using the balanced cantilever method. The model bridge was
built at half scale of this prototype and erected according to the balanced cantilever method
to approximately match the time- dependant stresses across the joints. A series of hydraulic
actuators and hollow- core jacks provided boundary conditions matching those on an
interior span of the prototype structure. The column used for the test was a hollow CIP
column with highly confined boundary elements. A similar column was also used in a
previous test and subjected an aggressive bi- directional seismic loading pattern [ 7].
The test was split into two stages. The first stage was designed to validate the existing
design philosophy of maintaining the superstructure of the bridge undamaged. To achieve
this the level of post- tensioning in the superstructure was selected to not allow any opening
of the segment- to- segment joints under plastic hinging of the column. The second stage
sought to allow some opening of the joints. For this purpose a portion of the superstructure
tendons were left unbonded in the initial stage and later removed. Additional vertical loads
simulating a vertical acceleration, were added to the structure resulting in a more severe
loading demand as well as a reduced capacity of the superstructure joints.
1.3 Issues To Be Addressed in Testing
The principle issues addressed by the segmental bridge test are listed below. A brief
explanation of the instrumentation and test setup designed to address each issue is also
listed.
5
• The potential for superstructure joints opening under seismic loading
o At least one linear potentiometer was placed across each of the joints and
several across the more critical joints to measure any opening across the
joints. Surface strain gauges were also placed near the joints to measure the
concrete strains.
• Different design objectives in the superstructure behavior under seismic loading
o In order to change the amount of prestressing present in the superstructure a
portion of the tendons were left unbonded to facilitate their removal during
the second stage of testing, and allow for different internal stress conditions.
• Any issues arising between the interaction of the column and superstructure
o A number of strain gauges were placed on the pier segment rebars that
could possibly undergo significant strains during testing. These gauges as
well as the gauges in the column were used to gather strain data in the pier
segment region of the specimen.
• The formation and performance of the plastic hinge in a hollow rectangular
column
o An extensive array of strain gauges was placed throughout the plastic hinge
region at the top of the column. Linear potentiometers were also used to
gather curvature information from the column. Three separate shear panels
were placed on the west side of the column to measure any shear
deformations of the column.
6
• Failure of the bridge system
o As well as the instrumentation on the test specimen clustered on the
northern and western faces of the unit, these sides were also used to mark
the formation and progression of cracks in the column and pier segment.
This allowed the other sides of the bridge to be clear for observation of
cracks and photographs with minimal obstructions. The strength and
stiffness of the system was measured to capture any deterioration during
testing.
1.4 Report Outline
The pages that follow discuss in detail the design, construction, analysis, and testing results
of the precast segmental bridge system test. A summary of each chapter follows.
Section 2: Test Unit Design and Details
In this chapter the selection and design of the prototype bridge is discussed and the tendon
layout considerations are presented. An overview of the design philosophy and detailing
of the half- scale test specimen constructed in the lab is also given.
Section 3: Construction and Material Properties
Chapter 3 presents the processes and challenges faced in construction of the test specimen.
An explanation of the precast segment construction process as well as the erection of the
column and placement of the segments in the lab is given. Material properties for all
concrete and steel reinforcing are tabulated.
7
Section 4: Test Setup, Instrumentation, and Loading Protocol
In this chapter the test setup, instrumentation, and loading procedures for both phases of
the segmental bridge test are explained. The test specimen was designed to be loaded in
the longitudinal direction under reversed- cyclic loading. The west and north faces of the
column were heavily instrumented, leaving the south and east faces for crack observation
and photographs.
Section 5: Analytical Considerations and Test Predictions
The analytical considerations of the testing program are presented in chapter 5. The
moment- curvature analysis performed and used to predict column deflection is outlined.
The post- tensioning loss estimates are presented. The predictions regarding the moments
and forces required to open the joints during both stages of the test are also given.
Section 6: Test Observations
Test observations are presented for each stage of the test. The formation and size of cracks
as witnessed during the test are explained. Additional photographs of the marked and
unmarked crack patterns are found in the photo appendix section. Cracks were marked at
the peaks of the first cycle at each ductility level. The cracks were marked with a blue pen
in the positive ( northward) push, and a red pen in the negative ( southward) direction.
Photographs and discussions of the joint opening and closing are also explained.
8
Section 7: Discussion of Test Results
Chapter 7 examines the measured results of each stage of the test. The hysteretic behavior
of the bridge system is presented and compared between the two stages of the test. System
displacements are examined and relevant strain gauge data is presented. Also the
measured joint behavior is presented in detail.
Section 8: Conclusions
This report discussed the design and testing of a half- scale bridge system built using the
balanced cantilever method of precast segmental construction. Issues regarding design and
analysis of precast segmental bridges are discussed based on the results from the segmental
bridge system test as well previous phases of the testing program [ 6]. Issues for future
research are highlighted where applicable.
9
2 Test Unit Design and Details
2.1 Overview
In this chapter the selection and design of the prototype bridge is discussed and the tendon
layout considerations are presented. An overview of the design philosophy and detailing
of the half- scale test specimen constructed in the lab is also given.
2.2 Prototype Bridge Structure
The prototype used throughout all phases of the segmental bridge test program was meant
to be a general bridge reflecting the types of situations in which a segmental bridge
construction scheme would be advantageous while still resulting in a test specimen that the
lab could accommodate. The prototype does not seek to model a specific existing bridge;
rather it was designed using characteristics generally associated with this type of bridge.
The prototype structure modeled in the bridge system test is a five- span segmental bridge
designed according to Caltrans seismic design criteria [ 3] with three interior 100’ ( 30.5m)
spans and exterior spans of 75’ ( 22.9m). The column height is 50’ ( 15.2m). The test
specimen models from mid- height of the column and to the midspan on each side of the
column ( Figure 2.1).
The bridge was designed with a segmental superstructure constructed using the balanced
cantilever method. This is different from the previous phases of the research project in
which a span- by- span construction scheme was used. However, for the systems test it was
decided that a balanced cantilever system would provide data more relevant to current
10
research needs. Although a 100’ span would be considered a very short span length for a
balanced cantilever bridge, lab space limitations and consistency with previous test phases
led to a decision not to lengthen the prototype bridge’s span length.
Figure 2.1 Prototype bridge structure
The superstructure box girder shape was based on the ASBI standard section for short
balanced cantilever span lengths [ 8]. The superstructure cross- section had a depth of 71 in.
( 1800 mm) and a total width of 331 in. ( 8400 mm). For the segments nearest to the
column a thicker bottom flange was needed due to the high negative moments during
construction and under seismic loading. Figure 2.2 shows the cross section of the main
prototype bridge box girder segment.
Figure 2.2 Prototype box girder cross section
100' ( 30.5m) 75' ( 22.9m)
50' ( 15.2m)
Bridge System Test Specimen
75' ( 22.9m) 100' ( 30.5m) 100' ( 30.5m)
450' ( 137.2m)
11
2.3 Prototype Column
The prototype substructure was hollow rectangular column with octagonal boundary
elements on each of the corners. These types of columns have been used recently in
several bridges in the San Francisco Bay Area [ 9,10]. High strength concrete ( 8000 psi)
was assumed in the prototype column.
The corner elements were sized to ensure that under a bi- directional seismic load ( Vs) and
plastic hinging of the column ( Mp) the confined portion of the corner would not crush ( see
Figure 2.3). The confined corner area’s compressive capacity ( Ccap) must be greater than
the axial load ( P) of the column plus resisting the compression force of all other corner
longitudinal bars yielding without crushing shown in equation 2.1.
Ccap ≥ Asl* No * fu + P ( 2.1)
where Asl is the area of the longitudinal steel and No is the number of longitudinal steel
bars excluding the bars in compression. The corner capacity is given by
Ccap = Acc* f'cc ( 2.2)
Where Acc is the confined corner area. Combining equations 2.1 and 2.2 gives the required
area of each confined corner:
cc
sl o u
cc f'
A A * N * f P
+
≥ ( 2.3)
12
Figure 2.3 Location of neutral axis in column under diagonal seismic loading
2.4 Prototype Tendon Layout
The layout of post- tensioning tendons in balanced cantilever construction involves two
steps. First the tendons are placed in the top flange of the superstructure to hold up the
segments during construction and resist the negative moment ( tension on top) of the
cantilevered arms. Then, once all the segments are in place the tendons are placed in the
bottom flange of the superstructure to resist positive moments.
Acc
Asl
Neutral
Compression Axis
Tension
P
Vs
Mp
P
13
Two separate design criteria were used to arrive at the desired tendon layouts for stages 1
and 2 of the test. For stage 1 the governing tendon condition was to limit all damage to the
column, thus a capacity design approach was used based on the ultimate moment ( Mu) of
the column. The seismic moment ( Ms) transferred to the superstructure was calculated as
follows.
2
)
2
( M V * h
M
s
u s
s
+
= ( 2.1)
Where Vs is the seismic shear in the column and hs is the height of the superstructure ( see
Figure 2.4).
Figure 2.4 Transfer of forces to superstructure during plastic hinging of column
Fs
Vs
Mu
Ms
Ms
hs
Fs
14
The tendon design for stage 2 sought to place the minimum number of strands in the
superstructure to achieve an inelastic response in the superstructure joints under severe
seismic loading. This design level was based on AASHTO strength design [ 1] and
allowing 3√ f’c tension under service loads.
Table 2- 1 summarizes the minimum number of 0.6” ( 15 mm) strands required across each
joint of the prototype for both design scenarios. Overall the reduction was approximately
25- 30% in the top joints and up to 70% in the bottom joints.
Table 2- 1 Minimum number of tendons required at each joint
Joint
Near Pier 2nd 3rd 4th 5th Midspan
Stage 1 Top 110 86 62 40 20 16
Bottom 46 58 66 68 70 72
Stage 2 Top 76 50 32 24 12 12
Bottom 12 18 24 36 44 46
2.5 Prototype Pier Segment
The pier segment of the prototype structure was initially designed according to standard
design principles [ 1,3,11]. One consideration was whether or not an access opening would
be needed through the pier segment. Due to the congestion of mild reinforcement and
tendon ducts in the limited area, the access opening was omitted under the assumption that
15
access to the interior of the box girder sections would be via an opening on the underside
of the bridge in each span.
Special attention was paid to the transfer of forces from the superstructure to the column
under opening of the first joint nearest the pier segment. Because the pier segment
extended past the face of the column additional reinforcement was needed to transfer shear
from the superstructure to the column. Once the joints open all the shear force would be
transferred through the compression region at the edge of the pier segment. This high
shear force was transferred via vertical headed bars to the column. Figure 2.5 shows the
transfer of the shear force Vb into the column using a strut- and- tie model in which the
compression struts are dashed lines and the tension ties solid lines. The tension of the
headed reinforcement is labeled Th.
Figure 2.5 Shear transfer through pier segment with opened joint
Fs
Vs
Mu
Ms
Vb
C C
C
T
Th
C
16
2.6 Test Unit Design and Details
The test specimen was built at half- scale of the prototype. Figure 2.6 shows the overall
layout of the test specimen. The superstructure segments are labeled 1 to 10 starting at the
south end of the bridge. The basic geometry and reinforcement configurations of the test
specimen are given in Figure 2.7 to Figure 2.14. The tendon layout geometry is shown in
Figure 2.15.
17
Figure 2.6 Test unit setup elevation view ( not to scale)
56' ( 17m)
78' ( 23.8m)
nose
Reaction frame
CIP end
block
Steel
( Extra DL)
CIP
Column
Loading beams
for additional vertical loads
31' ( 9.4m) 31' ( 9.4m) 8'
PT bar
8'
Reaction wall
Not to
Scale
Pin at column
midheight
CIP
Diaphragm
16' ( 4.9m)
CIP closure
joints
18
Figure 2.7 Cross section of test unit column plastic hinge region
Figure 2.8 Cross section of test unit column non- plastic hinge region
19
Figure 2.9 Elevation of test unit column and footing
20
Figure 2.10 Test unit pier segment reinforcement detail
21
Figure 2.11 Cross section of test unit segments 5 and 6
Figure 2.12 Cross section of test unit segments 1- 4 and 7- 10
22
Figure 2.13 Test unit footing reinforcement detail
23
Figure 2.14 Test unit end block reinforcement detail
24
Figure 2.15 Test unit tendon layout
2.7 Strand Bond Area Scaling
Matching the bond area as closely as possible in the scaled down model was important to
accurately portray the debonded length of the grouted segments. It was found that by using
0.5” strands in the test unit compared to the 0.6” strands in the prototype resulted in very
close bond area matches of the bonding area necessary in the key tendons nearest the
column [ 12].
The bond area per unit length ( U) was calculated according to Eqn 2.4 ( 8).
U = π * ds * ns * rs * N* aave
( 2.4)
25
Where ds is the diameter a strand, ns is the number of exterior wires on each strand, rs is the
ratio of the exterior portion of each wire to the interior portion ( assumed to be 220/ 360), N
is the number of exterior strands, and aave is the average number of exposed sides of the
exterior strands.
26
3 Construction and Material Properties
3.1 Overview
Chapter 3 presents the processes and challenges faced in construction of the test specimen.
An explanation of the precast segment construction process as well as the erection of the
column and placement of the segments in the lab is given. Material properties for all
concrete and steel reinforcing are tabulated.
3.2 Construction of the Bridge System Test Unit
A series of additional illustrative photographs of the construction are found in the appendix
of this report. The scale and precarious nature of the test unit made its construction an
interesting and challenge task. To begin with the footing was cast and moved into position
on the lab floor. After this each spiral corner of the column was tied separately and lifted
into position on the footing. A foam blockout was used to maintain the gap between the
footing and the column base. All the longitudinal steel of the column was placed and the
bottom 30 in., the solid section, was poured. Following this the remainder of the column
transverse steel and stirrups were placed and the middle and top sections of the column
were cast in two separate pours. The same set of forms was used for all three pours,
moved up and fastened higher on the column each time. A platform was then constructed
around the column top upon which the forms and steel for the pier segment were erected.
Careful consideration was given to the location of all post- tensioning ducts to insure that
the tendons would be able to pass through the section once all the segments were in place.
27
While the construction of the column and pier segment was going on at the Powell Labs in
San Diego, California; the individual segments were cast by Keicon Precasters located in
Stockton, California. The ten precast segments were cast using a single steel form with
interchangeable interior pieces to accommodate the different interior dimensions and
blister details. The pieces were match- cast starting one of the segments closest to the
column and then proceeding along one side. After the five on one side of the column were
finished those on the opposite column side would be cast in a similar manner. Once begun,
the casting of all ten precast segments was completed in ten working days.
Once the segments were delivered to the lab, platforms were raised on either side of the
column and the first pair of segments positioned atop them leaving space for the three inch
closure gap. The closure gap was poured and allowed to set for two weeks before the
permanent cantilever tendons for the first pair of segments were post- tensioned. The
process of post- tensioning the sequential pairs of segments was then repeated as shown in
Figure 3.1. First the platforms moved out from the previous pair, and the new segments
were lifted into place and segmental bridge adhesive epoxy was applied to each face to be
joined ( Step A). Next, four temporary post- tensioning bars were placed through the bridge
structure to provide an even pressure across the faces during curing of the epoxy, and to
hold the segments in place before the permanent tendons were installed ( Step B). After the
permanent tendons were installed ( Step C) the temporary bars were removed and the
process repeated until all ten segments ( five pairs) were erected.
The end blocks used to apply the boundary conditions were then placed and a closure joint
was poured between the blocks and the outermost pair of segments of the test specimen.
28
With the blocks in place it was then possible to install the tendons along the bottom of the
superstructure as well as the continuity tendons and unbonded tendons along the top.
After all tendons were installed in the bridge, the fixtures were attached to the bridge. The
wiffle tree and steel nose systems were installed and all eight actuators hooked up to allow
the application of loads and boundary conditions.
Figure 3.1 Erection sequence of precast segments
3.3 Material Properties
The following section outlines the material properties for concrete and steel used in the
test. Design concrete strength was f’c = 8 ksi ( 55 MPa) for the column and pier segment
and 5 ksi ( 35MPa) for the segments.
Permanent
Tendon
Temporary
PT bars
Segment
awaiting
placement
Previously
erected segments
Stage A Stage B Stage C
29
3.3.1 Concrete Properties
The column was poured in five stages. The footing block first, followed by the solid pin
section which made up the bottom 30 inches ( 762mm) of the column. The hollow portion
of the column was poured in two separate lifts called column middle and column top.
Lastly the pier segment was poured on top of the column. The 3” closure gaps on either
side of the pier segment and those connecting the end blocks to the precast segments used
the same concrete mix design as the column and pier segment.
In order to insure the concrete was properly distributed throughout the closely spaced
reinforcement of the corner elements and walls a self- consolidating concrete was used.
This required practically no vibration of the concrete resulting in limited aggregate
separation, and ease of placing the concrete. A minimal spread of 25 inches was specified
for the self- consolidating concrete.
The compressive strengths were determined according to the average of three 12 in. tall by
6 in. diameter cylinder tests according to standard practice. The day- of- test ( DOT)
strengths were taken during the first two days of testing. Because the critical concrete pour
sections of the column top, pier segment, and pier segment closure gap were old at the
beginning of testing ( at least 91 days), no significant differences in strength were
anticipated over the weeks of testing. A few smaller cylinders of the grout mix were also
tested. The DOT strength of the segments was taken as the average of two cylinders
shipped with each segment. All post- test analysis were done using the average DOT
strength for each component. Table 3- 1 and Table 3- 2 give the results of the compressive
30
strength tests for all sections. Split cylinder tests were performed on the column top to
gauge the DOT tensile strength. The result was a splitting stress of 0.64 ksi ( 4.4 MPa).
Table 3- 1 Test unit concrete compressive cylinder strengths ( ksi)
Date Cast 7- day
14-
day
28 day DOT
Age on
DOT
( days)
Column Bottom 11- 13- 03 12.1 201
Column Middle 12- 5- 03 11.5 179
Column Top 1- 13- 04 11.6 140
Pier Segment 2- 24- 04 7.7 9.0 10.0 11.6 98
Pier Gap 3- 2- 04 7.8 10.3 12.5 91
End Gap 4- 27- 04 8.1 10.2 35
Grout 5- 10- 04 4.3 31
Table 3- 2 Test unit segment compressive day of test strengths ( ksi)
Segment 1 2 3 4 5 6 7 8 9 10
DOT
Strength
8.8 9.3 9.7 9.1 9.2 9.0 7.8 8.9 8.2 5.9
31
Table 3- 3 Test unit concrete compressive cylinder strengths ( MPa)
Date Cast 7- day
14-
day
28
day
Day of
Test
Age on
DOT
( days)
Column Bottom 11- 13- 03 83 201
Column Middle 12- 5- 03 79 179
Column Top 1- 13- 04 80 140
Pier Segment 2- 24- 04 53 62 69 80 98
Pier Gap 3- 2- 04 54 71 86 91
End Gap 4- 27- 04 56 70 35
Grout 5- 10- 04 30 31
Table 3- 4 Test unit segment compressive day of test strengths ( MPa)
Segment 1 2 3 4 5 6 7 8 9 10
DOT Strength 61 64 67 63 63 62 54 61 57 41
3.3.2 Steel Properties
Table 3- 5 and Table 3- 6 give the properties for the mild reinforcement used in the test
specimen. The reinforcement used in the footing, the segments, and the end blocks is not
included. The numbers given are the averages of three separate bars of each size taken
from each batch of steel used. The high- strength strands used were 0.5” ( 13 mm) diameter,
low relaxation strands. Since no yielding of the strands was anticipated and due to the
difficulty in producing reliable material test results from the 7- wire strands, no material
tests were performed. In case of unexpected yielding, strand sections from the main spool
were retained for possible future testing.
32
Table 3- 5 Test unit mild steel reinforcement properties ( ksi)
Fy Fu
# 3 bar – Column. Long wall transverse bar 65.9 103.8
# 3 bar – Column. Short wall transverse bar 65.6 99.2
# 4 bar – Column, Longitudinal group A 64.8 105.7
# 4 bar – Column. Longitudinal group B 65.1 105.0
# 4 bar – Pier Segment 66.5 87.8
Table 3- 6 Test unit mild steel reinforcement properties ( MPa)
Fy Fu
# 3 bar – Column. Long wall transverse bar 455 716
# 3 bar – Column. Short wall transverse bar 453 684
# 4 bar – Column, Longitudinal group A 447 729
# 4 bar – Column. Longitudinal group B 449 725
# 4 bar – Pier Segment 459 606
33
4 Test Setup and Instrumentation
4.1 Overview
In this chapter the test setup, instrumentation, and loading procedures for both phases of
the segmental bridge test are explained. The test specimen was designed to be loaded in
the longitudinal direction under reversed- cyclic loading. The west and north faces of the
column were heavily instrumented, leaving the south and east faces for crack observation
and photographs.
4.2 Test Setup
The test unit was loading in the longitudinal direction according to a displacement- based,
incrementally- increased, fully- reversed cyclic loading pattern. The lateral load was
applied via four +/- 225 kip, 48 in. ( 1000kN, 1.2m) stroke MTS actuators. The two
northern horizontal load actuators were anchored to the lab strong wall and the southern
horizontal actuators were held in place by steel a- frames atop concrete blocks. Two 300
kip ( 1320kN) hollow core jacks provided the gravity load. Each jack was attached to a
high strength steel bar and applied the force to a series of steel beams called a wiffle tree.
These wiffle trees spreads the single point load on either side of the span to 64 points
across the deck of the bridge structure better simulating a distributed load. The axial load
ratio of the column during stage 1 was 0.027 and 0.047 during stage 2. Because the bridge
system represents a single span of a continuous span bridge the moments at the ends of the
spans are not zero. A pair of vertical actuators on each end applied a constant moment at
the bridge ends. These actuators were +/- 225 kip, 18 in. ( 1000kN, 0.46m) stroke MTS
actuators. One applied a force directly under the end block and the other and equal and
34
opposite force with a moment arm of 8ft via the steel nose pieces attached to the bridge
ends. Figure 4.1 and Figure 4.2 show the scale schematics of the test set- up and a photo is
shown in Figure 4.3.
Figure 4.1 Test unit east- west elevation view
Figure 4.2 Test unit plan view
35
Figure 4.3 Test unit photo
4.3 Loading History
The test unit was subjected to a reversed cyclic loading pattern in the longitudinal
direction. The loading history is shown in Figure 4.4. Two cycles at increasing ductility
levels were used in order to allow for further testing of the system without considerable
strength deterioration in the column. Testing stage 1 consisted of the cycles up to and
including a system displacement ductility of 4 ( μ= 4) which is a minimum requirement for
seismically designed bridges. During stage 1 the vertical load and end moments applied to
the superstructure were equivalent to those caused only by gravity loads.
36
At the commencement of stage 2 a cycle was repeated at μ= 4 to easily compare any
differences in behavior between the two stages. During stage 2 the vertical load and end
moments applied to the superstructure were equivalent to a vertical acceleration of 0.75g
plus the gravity load.
Figure 4.4 Loading history
4.4 Instrumentation
4.4.1 Mild Reinforcement Strain Gages
The strain of reinforcing bars in the column and pier segment were monitored using strain
gauges. Most gauges were 12Ω resistance gages with a 5mm ( 0.2 in.) gauge length. In a
- 8
- 6
- 4
- 2
0
2
4
6
8
Stage 1
Stage 2
37
few locations near the column- pier segment interface, post- yield gauges were used to
capture post- yield strains in the bars. The reinforcing bars were prepared by grinding
smooth a section of bar, roughening the surface, and cleaning it with methyl ethyl- keytone.
The gauges were then applied to the prepared surface with an adhesive ( alpha
cyanoacrylate monomer). To protect them from water and mechanical abrasions an acrylic
based water- proofing agent and a vinyl mastic membrane were applied over the gauges.
Figure 4.5 and Figure 4.6 show the strain gauge details of the bridge column. The
longitudinal bar gauge locations were selected to give a strain profile of the column. On
the transverse bars gauges were added to measure overall strain in the bar as well as gain
information on development of the straight bars ( 13). The strain gauge layout of the pier
segment is shown in Figure 4.7. These gauges were intended to monitor the transition of
forces through the pier segment.
Figure 4.5 Column reinforcement strain gauge layout
38
Figure 4.6 Column reinforcement strain gauge elevation view
Figure 4.7 Pier segment reinforcement strain gauge layout
39
4.4.2 Tendon Strain Gauges
Numerous strain gauges were attached to the high strength steel strands in a similar manner
to the mild reinforcement gauges. There was a very high failure rate of these gauges
during insertion of the strands, stressing, and grouting of the tendons. After stressing and
grouting of the tendons only a third of the gauges on the tendons survived. Due to these
difficulties, which increased with the lengths of the tendons gauged, limited reliable test
data was gleaned from these gauges. All the gauges that survived the rigors of stressing
and grouting are shown in Figure 4.8.
Figure 4.8 Tendon strain gauge layout
4.4.3 Concrete Strain Gauges
Surface gauges were used to monitor the strain along the top and bottom of the
superstructure. Figure 4.9 shows the layout of these gauges on either face of the bridge
deck. The gauges were placed close to the superstructure joint so that upon opening an
easily detectable difference between the strain gauges and the linear potentiometers
crossing the joints would exist.
40
Figure 4.9 Concrete gauge layout
4.4.4 Curvature Instrumentation
Along the length of the column’s west side displacement potentiometers were placed to
calculate curvature of the column. This was done according to Eqn. 4.1
g
n s
D * L
Δ Δ
φ
φ
−
= ( 4.1)
Where Φ is the average curvature at the elevation of the devices along the gauge length, Δn
and Δs are the potentiometer readings on the north and south sides, DΦ is the distance
between the potentiometers, and Lg is the gauge length. Curvatures were calculated as
positive in the push ( north) direction and negative in the pull ( south) direction. Figure 4.10
shows the layout of curvature instrumentation. To avoid disturbances from crushing and
spalling of cover concrete the gauges were anchored inside the corner spiral element
reinforcing.
41
At the boundary of the column and pier segment a single crack was expected, therefore the
topmost instrument measure the rotation of the column relative to the pier segment. This
rotation ( θ) was calculated as
Dφ
θ Δ Δ n − s
= ( 4.2)
Figure 4.10 Column curvature instrumentation layout
4.4.5 Shear Instrumentation
The measurement of shear deformation in the column was achieved by two independent
linear potentiometers arranged diagonally on the west face forming a shear panel.
42
Under flexure and expansion of the panel region the diagonal deformations remain equal as
shown in Figure 4.11 and Figure 4.12. The following figure ( Figure 4.13) shows that only
in shear do the diagonal deformations differ.
D*
h
d2
d1
d1’
d2’
Figure 4.11 Equivalent diagonal deformations under flexure
43
D*
h
d2
d1
d1’
d2’
Figure 4.12 Equivalent diagonal deformations under horizontal and vertical expansion
D*
h d2
d1
d1’
δ2 d2’
Δ2
γ2
δ1
Δ1
γ1
Figure 4.13 Diagonal deformations used to calculate shear deformation
44
Using these diagonal displacement measurements and assuming small angles, the average
shear deformation was calculated according to equation 4.3.
2h
Δ Δ
h
γ γ γ 1 2 1 + 2
=
+
= ( 4.3)
The lateral deformations Δ1 and Δ2 due to shear deformation are calculated from the
diagonal deformations using the ratio
D*
d
δ
Δ = ( 4.4)
where
δ = d'− d ( 4.5)
Combining equations 4.3 and 4.4 yields the equation
2hD*
γ δ d δ d 1 1 − 2 2
= ( 4.6)
This equation gives the average shear deformation over a given region with height h, and
depth D*. Figure 4.14 shows the shear panel instrumentation on the west face of the
column. The panels where attached to the column corners by imbedded 3/ 8 in. ( 9.5mm)
anchors.
45
Figure 4.14 Shear instrumentation layout
4.4.6 Linear Potentiometers
Linear displacement devices were used in a variety of locations on the test specimen to
monitor its performance. In order to have a record of the deformations along the length of
the superstructure vertical displacements of the superstructure were monitored by string
potentiometers at each end, the midspans, and center of the bridge. Across most of the
superstructure joints displacements were measured on the top and some along the bottom
as well. The locations of these potentiometers are shown in Figure 4.15. Linear devices
were also used to monitor rotation at the column pin as well as any uplift in the footing.
46
Figure 4.15 Joint displacement potentiometer layout
4.4.7 Inclinometers
Inclinometers were placed on the bridge specimen to measure rotations and corroborate
information gained from linear displacement devices. The inclinometers were placed at
each end block, at the base of the column, and on the pier segment as shown in Figure
4.16.
Figure 4.16 Inclinometer device layout
47
5 Analysis and Predictions
5.1 Overview
The analytical considerations of the testing program are presented in chapter 5. The
moment- curvature analysis performed and used to predict column deflection is outlined.
The post- tensioning loss estimates are presented. The predictions regarding the moments
and forces required to open the joints during both stages of the test are also given.
5.2 Moment- Curvature Analysis
Using the program ANDRIANNA, moment curvature analysis were performed for the
column and superstructure using non- linear material models for the concrete ( both
confined and unconfined), the mild reinforcement, and the prestressing steel in the
superstructure.
As outlined in the program manual [ 14] the concrete material model used in the program
for both confined and unconfined concrete uses a stress- strain curve based on Mander’s
Model [ 15]. The program allows for tensile cracking as well as compressive strain
hardening and softening. The mild steel reinforcement is linear elastic up until reaching
the yield stress, at which point a perfectly plastic yield plateau occurs in the stress- strain
diagram. Strain hardening occurs after the yield plateau, curving parabolically until the
ultimate stress is reached. The prestressing steel uses the model by Collins and Mitchell
[ 16], based on the Minegotto- Pinto function [ 17]. Figure 5.1 to Figure 5.3 show the
material models used. Figure 5.4 and Figure 5.5 show the results of the moment curvature
48
analysis for the column under dead load and the analysis for the superstructure segments
nearest the column.
0.000 0.005 0.010 0.015 0.020 0.025
Strain
0
2
4
6
8
10
12
14
16
Stress ( ksi)
0
20
40
60
80
100
Stress ( MPa)
Figure 5.1 Concrete material model
0.00 0.02 0.04 0.06 0.08 0.10 0.12 0.14
Strain
0
20
40
60
80
100
120
Stress ( ksi)
0
100
200
300
400
500
600
700
800
Stress ( MPa)
Figure 5.2 Mild steel material model
49
0.00 0.01 0.02 0.03 0.04 0.05
Strain
0
40
80
120
160
200
240
280
Stress ( ksi)
0
200
400
600
800
1000
1200
1400
1600
1800
Stress ( MPa)
Figure 5.3 Prestressing steel material model
0.0000 0.0005 0.0010 0.0015 0.0020 0.0025
Curvature ( rads/ in)
0
50000
100000
150000
200000
Moment ( k- in)
0
5000
10000
15000
20000
Moment ( kN- m)
0.00 0.01 0.02 0.03 0.04 0.05 0.06 0.07 0.08 0.09
Curvature ( rads/ m)
Figure 5.4 Column moment- curvature results
50
- 0.0004 - 0.0002 0.0000 0.0002 0.0004 0.0006
Curvature ( rads/ in)
- 400000
- 300000
- 200000
- 100000
0
100000
200000
Moment ( k- in)
- 50000
- 40000
- 30000
- 20000
- 10000
0
10000
20000
Moment ( kN- m)
- 0.011 - 0.006 - 0.001 0.004 0.009 0.014 0.019 0.024
Curvature ( rads/ m)
Figure 5.5 Superstructure moment- curvature results
5.3 Column Deflection
The total deflection of the column ( Δt) was predicted as the sum of the elastic ( Δe) and
plastic ( Δp) displacement components.
Δt = Δe + Δp ( 5.1)
With
3M '
Δ ' L * M
y
2
y
e
Φ
= ( 5.2)
51
) L * L
M '
Δ ( Φ Φ ' M p
y
p = − y ( 5.3)
Where M is the moment at a given displacement, Φ is the curvature at that displacement,
Φy’ is the theoretical yield curvature, My’ is the theoretical yield moment, L is the column
shear span, and Lp is the column plastic hinge length.
The plastic hinge length ( Lp) is calculated according to Eqn. 5.4 [ 10].
Lp = 0.08L + αD* + 9db ( 5.4)
Where D* is the confined concrete section depth, db is the diameter of reinforcing bar, and
the constant α is assumed to be 0.30. The force- displacement predictions for the test unit
are shown in Figure 5.6.
52
- 10 - 8 - 6 - 4 - 2 0 2 4 6 8 10
- 200
- 150
- 100
- 50
0
50
100
150
200
250
- 5 - 4 - 3 - 2 - 1 0 1 2 3 4 5
Drift ratio (%)
- 9 - 8 - 7 - 6 - 5 - 4 - 3 - 2 - 1 0 1 2 3 4 5 6 7 8 9
Stage 1
Prediction
Stage 2
Prediction
- 800
- 600
- 400
- 200
0
200
400
600
800
1000
Lateral load ( kN)
Displacement at top of the column ( in.)
Lateral load ( kips)
Displacement ductility factor
Figure 5.6 Force- displacement prediction for test unit
5.4 Prestress Losses
The anticipated losses in the prestress tendons was calculated according to AASHTO
guidelines [ 1,2] as well as more detailed methods [ 18] as outlined below for elastic
shortening, friction, anchorage seating, and approximated long term losses. It was
assumed in the prototype structure that after all losses the tendons would be stressed at
0.60* fpu or 162 ksi ( 1118 Mpa). The jacking stress of each tendon in the specimen took
into account the losses as outlined below to achieve a stress level of approximately 60% of
the ultimate stress. A summary of the tendon losses is given in Table 5- 1.
53
5.4.1 Elastic Shortening Losses
The elastic shortening loss ( ΔfpES) was calculated according to Eqn. 5.5
ΔfpES = ( n * f )
n
1 n
j 1
j
j
csj Σ=
( 5.5)
Where nj is the number of subsequently stressed tendons and the stress in the concrete, at
the location each tendon due to the subsequent tendons is fcsj.
5.4.2 Friction Losses
The friction loss ( ΔfpF) was calculated according to Eqn. 5.6.
ΔfpF = fpj * ( α* μ+ K* L) ( 5.6)
Where fpj is the jacking stress; α is angle change of the tendon; μ, the coefficient of
friction, was assumed to be 0.25; and K is the wobble coefficient ( 0.0002 / ft).
5.4.3 Anchorage Seating Losses
The anchorage seating loss ( ΔfpA) was calculated according to Eqn. 5.7
54
ΔfpA =
L
E* a ( 5.7)
Where a, the expected seating length, was assumed to be 0.25 in.
5.4.4 Long Term Losses
The long term losses estimated for the test specimen come from contributions of creep
( ΔfpCR), shrinkage ( ΔfpSH), and relaxation ( ΔfpR).
ΔfpLT = ΔfpCR + ΔfpSH + ΔfpR ( 5.8)
The creep loss ( ΔfpCR) was calculated according to Eqn. 5.9
ΔfpCR = K ( f f )
E
E
cr cs csd
c
s − ( 5.9)
Where Kcr is the 1.6, fcs is the stress in the concrete at the level of prestressing immediately
after transfer, and fcsd is the stress at the same location due to all superimposed dead loads
applied after the post- tensioning is applied.
The shrinkage loss ( ΔfpSH) was calculated according to Eqn. 5.10
ΔfpSH = 0.8( 17,000 – 150* RH) ( 5.10)
Where RH is the relative humidity in percent ( assumed to be 70%)
The relaxation loss ( ΔfpR) was calculated according to Eqn. 5.11
55
ΔfpR = 5 – 0.07 ΔfpF – 0.1 ΔfpES – 0.05( ΔfpSH + ΔfpCR) ( 5.11)
Table 5- 1 Specimen theoretical prestress losses in ksi ( MPa)
Tendon
Friction
( ΔfpF)
Anchorage
( ΔfpA)
Elastic
Shortening
( ΔfpES)
Long- Term
Losses
( ΔfpLT)
Total
( ΔfpTot)
A 28 12 0 15 55
B 19 19 0 15 58
C 6 45 2 14 67
D 10 26 1 14 51
E 24 14 0 15 53
F 16 84 0 10 110
G 26 22 0 9 58
H 25 10 3 11 49
J 26 10 3 17 56
5.5 Joint Opening
An initial prediction was made as the whether the joints were likely to open during the test.
The anticipated moment at the joint locations from both the dead load and seismic loads
transferred from plastic hinging of the column was compared to the theoretical cracking
moment at the joint. Figure 5.7 and Figure 5.8 show how this theoretical moment demand
at the joint locations compared to the decompression moment as well as a cracking
moment when a tensile strength of 6√ f’c is assumed. . Figure 5.7 shows that the moment
range expected in the test ( the blue line) stays within the bounds of the decompression
56
moments meaning that no tensile strains were expected during the first stage of testing. As
shown in Figure 5.8 where the moment demand curve meets the cracking capacity curve,
the joints nearest to the pier were expected to undergo some opening during the second
stage of testing,
Figure 5.7 Joint opening prediction for stage 1
- 6 - 5 - 4 - 3 - 2 - 1 0 1 2 3 4 5 6 7
Joint from Column
- 2500
- 2000
- 1500
- 1000
- 500
0
500
1000
1500
2000
Moment ( k- ft)
- 4000
- 3000
- 2000
- 1000
0
1000
2000
Moment ( kN- m)
Stage 1 Loads
Top Decompression
Top Cracking
Bottom Cracking
Bottom Decompression
57
- 6 - 5 - 4 - 3 - 2 - 1 0 1 2 3 4 5 6 7
Joint from Column
- 2500
- 2000
- 1500
- 1000
- 500
0
500
1000
1500
2000
Moment ( k- ft)
- 4000
- 3000
- 2000
- 1000
0
1000
2000
Moment ( kN- m)
Stage 2 Loading
Top Cracking
Bottom Cracking
Top Decompression
Bottom Decompression
Figure 5.8 Joint opening prediction for stage 2
58
6 Test Observations
6.1 Overview
Test observations are presented for each stage of the test. The formation and size of cracks
as witnessed during the test are explained. Additional photographs of the marked and
unmarked crack patterns are found in the photo appendix section. Cracks were marked at
the peaks of the first cycle at each ductility level. The cracks were marked with a blue pen
in the positive ( northward) push, and a red pen in the negative ( southward) direction.
Photographs and discussions of the joint opening and closing are also explained.
6.2 Test Stage 1
Testing of the bridge structure began on June 10th. After the vertical loads and boundary
condition moments were applied to the structure, cyclic loading began. The first loading
peak reached was at a displacement equivalent to half the theoretical yield displacement
( 0.5Δy). At this level some hairline flexure cracks appeared in the column ( Figure 6.1).
After yield displacement ( Δy) was reached the cracks were imperceptible when the seismic
loads were removed from the bridge. At ductility level 2 ( μ= 2) the flexure cracks on the
tension side spread down the column. Shear cracks also formed on the east and west faces
of the column ( Figure 6.2). The largest crack width measured at μ= 2 was a 0.4 mm tension
crack near the top of the column. During the early cycles of testing the bridge exhibited a
propensity to twist likely due to some construction irregularities likely in the column. The
actuators were programmed to inhibit this twist and the applied torque from the actuators
was closely monitored. After the completion of stage1 testing the actuator control program
was changed to better control the twisting while limiting the torque forces put into the
59
system. At μ= 4 there was significant flexural cracking throughout the column shown with
and without marked cracks ( Figure 6.3 and Figure 6.4). There was not yet any spalling of
the concrete at μ= 4. The initiation of short vertical cracks running between the large
tensile flexural cracks near the top on the corners signified the initiation of spalling, but
none had yet occurred. This was significantly different from the testing of a similar
column in a severe bi- directional loading path [ 7] in which spalling began before μ= 4 was
reached.
Up to the end of testing stage 1 no significant cracking was evident in the superstructure
joints. At the highest ductility level of stage 1 a slight hairline crack was seen in the first
joint opening. This was likely the reopening of a creep or shrinkage crack in the closure
pour region. When the seismic loads were removed the hairline crack was no longer
visible.
60
Figure 6.1 Marked crack pattern on north and west faces at initial displacement level of 0.5Δy
Figure 6.2 Marked crack pattern on north and west faces of column at μ= 2
61
Figure 6.3 Marked crack pattern on north and west faces of column at μ= 4
Figure 6.4 Unmarked crack pattern on south and east faces of column at μ= 4
62
6.3 Test Stage 2
The second stage of testing began on August 17th. The first cycle was a repeat of the μ= 4
cycle at the end of stage 1. During stage 2 openings in the superstructure joints nearest the
column were evident under the same ductility level achieved previously ( μ= 4). The total
crack width at the joint was measured to be approximately 0.4 to 0.5 mm. This was the
combined crack widths of each side of the 3 inch closure gap. A close up of this crack at
the joint is shown in Figure 6.5. The maximum crack width in the column was measured
as between 2.5 and 3 mm. At μ= 5 extensive spalling of the concrete in the corner regions
began ( Figure 6.6), and the superstructure joint crack was measured at approximately 0.6
mm. At μ= 6 the joint crack was measured at around 0.8 mm and in subsequent cycles and
increased ductility levels was not seen to exceed this level by much. No opening of any
joints other than those nearest the column was observed. Through the entire loading
pattern upon removing the seismic loads the superstructure cracks would close. At
ductility levels of μ= 6 and μ= 8, the column had severe flexural cracking present ( Figure
6.7 and Figure 6.8). In the final cycles of the test heavy spalling occurred at all corners of
the column. No spalling occurred in the wall section. The test was concluded after two
cycles at μ= 8 because of concerns of stability of the bridge specimen if it was more heavily
damaged. Deterioration of the column would have continued with more cycles and higher
ductility levels, leading to failure of the column.
63
Figure 6.5 Top of joint crack close up during stage 2 at μ= 4
64
Figure 6.6 Spalling of column south face corner regions at μ= 5
Figure 6.7 Marked crack pattern on north and west faces of column at μ= 6
65
Figure 6.8 Marked crack pattern on north and west faces of column at μ= 8
66
7 Discussion of Test Results
7.1 Overview
Chapter 7 examines the measured results of each stage of the test. The hysteretic behavior
of the bridge system is presented and compared between the two stages of the test. System
displacements are examined and relevant strain gauge data is presented. Also the
measured joint behavior is presented in detail.
7.2 Test Stage 1
7.2.1 Force- Displacement Behavior
The force- displacement data for stage 1 of the test unit, along with the initial prediction is
shown in Figure 7.1. The overall performance was typical of a well- confined, reinforced
concrete column showing high- energy dissipation capacity, high displacement ductility.
The response at each ductility level was similar during the first and second cycles. The
earlier monotonic prediction for the stage 1 is also shown on the graph. The prediction
underestimated the strength of the column by about 7%, likely due to variations in material
properties and construction irregularities. Some of the irregular loops and glitches near the
zero load mark were due to controlling the test specimen and bringing it back to a stable
position at the end of each day’s testing. The shear displacements were insignificant
compared to the flexural displacements according to the measurements taken.
67
- 10 - 8 - 6 - 4 - 2 0 2 4 6 8 10
- 250
- 200
- 150
- 100
- 50
0
50
100
150
200
250
- 5 - 4 - 3 - 2 - 1 0 1 2 3 4 5
Drift ratio (%)
- 9 - 8 - 7 - 6 - 5 - 4 - 3 - 2 - 1 0 1 2 3 4 5 6 7 8 9
Stage 1
Prediction
- 1000
- 600
- 200
200
600
1000
Lateral load ( kN)
- 200 - 100 0 ( mm) 100 200
Displacement at top of the column ( in.)
Lateral load ( kips)
Displacement ductility factor
Stage 1
Figure 7.1 Stage 1 force- displacement diagram
7.2.2 Curvature
The curvature profile of the column for stage 1 of the test is shown in Figure 7.2. With the
height measured from the base of the column, the formation of the plastic hinge zone at the
top of the column is evident. The theoretical yield curvature ( Φy) is shown on the positive
and negative sides of the graph. In the higher ductility levels the curvature significantly
exceed Φy at a height of approximately 118 inches giving a plastic hinge length of around
30 in ( 760mm).
68
- 0.0006 - 0.0004 - 0.0002 0.0000 0.0002 0.0004 0.0006 0.0008 0.0010
Curvature ( rads/ in.)
0
24
48
72
96
120
144
Height above footing ( in.)
0
500
1000
1500
2000
2500
3000
3500
Height above footing ( mm)
- 0.02 - 0.01 0.00 0.01 0.02 0.03
Curvature ( rads/ m)
Ductility = 1
Ductility = 2
Ductility = 3
Ductility = 4
Theoretical
yield curvature
Figure 7.2 Stage 1 curvature profile
7.2.3 Longitudinal Bar Behavior
Several of the longitudinal bars in the column were instrumented to capture the strain
profile of the bars. Figure 7.3 shows the locations of the gauged longitudinal bars. The
tensile strains along the lengths of these bars at varying ductility levels are shown in Figure
7.4. The gaps in the graph arise from damaged gauges. At increasing displacement levels
more gauges were damaged resulting in more gaps. Using a theoretical yield strain of
0.005 ( 5000 microstrain) the spread of the plastic hinge down the column is evident. At
μ= 4 the apparent plastic hinge length is around 30 in. ( 760mm). This corroborates with the
69
data from the curvature instrumentation. The strain penetration of the longitudinal bars into
the pier segment appears to be around 12 in. ( 305mm) at μ= 4.
Figure 7.3 Location of gauged longitudinal bars
0 5000 10000 15000 20000
Microstrain ( in/ in or mm/ mm)
80
100
120
140
160
180
Height above footing ( in.)
2000
3000
4000
5000
( mm)
Bar SLS2A
Column Top
0 5000 10000 15000 20000
Microstrain ( in/ in or mm/ mm)
80
100
120
140
160
180 Height above footing ( in.)
2000
3000
4000
5000
( mm)
Bar SLS2F
Column Top
70
0 5000 10000 15000 20000
Microstrain ( in/ in or mm/ mm)
80
100
120
140
160
180
Height above footing ( in.)
2000
3000
4000
5000
( mm)
Bar SLS3C
Column Top
0 5000 10000 15000 20000
Microstrain ( in/ in or mm/ mm)
80
100
120
140
160
180
Height above footing ( in.)
2000
3000
4000
5000
( mm)
Bar SLS6C
Column Top
0 5000 10000 15000 20000
Microstrain ( in/ in or mm/ mm)
80
100
120
140
160
180
Height above footing ( in.)
2000
3000
4000
5000
( mm)
Bar SLS8A
Column Top
0 5000 10000 15000 20000
Microstrain ( in/ in or mm/ mm)
80
100
120
140
160
180
Height above footing ( in.)
2000
3000
4000
5000
( mm)
Bar SLS8F
Column Top
Figure 7.4 Longitudinal bar strain profiles for stage 1 at increasing ductility levels
7.2.4 Transverse Bar Behavior
The transverse bars in the short wall of the column were gauged to measure shear along the
length of the column. The locations of the gauged bars are shown in Figure 7.5. At the top
and bottom of the column the strain path is irregular due to cracking and the transfer of the
forces into the pin respectively. However along most of the column length the strains in
the transverse reinforcement were fairly constant. This can be seen in Figure 7.6. The
strain in the bars did not exceed the theoretical yield strain ( εy). Some inherent
irregularities due to construction and possibly material differences led to a propensity
71
towards some twisting of the column during testing. Because the base of the column was a
pin the loading actuators solely inhibited the twisting in order to keep the bridge structure
loading in one direction. Because of this a small toque force was seen in the column and
some differences were seen in the column strain reading for the push and pull directions.
This data is presented in the appendix.
Figure 7.5 Location of gauged transverse bar
300 800 1300 1800
Microstrain
20
40
60
80
100
120
140
160
Height above footing ( in.)
SRT5A
1000
1500
2000
2500
3000
3500
4000
Height above footing ( mm)
300 800 1300 1800
Microstrain
20
40
60
80
100
120
140
160
Height above footing ( in.)
SRT5F
1000
1500
2000
2500
3000
3500
4000
Height above footing ( mm)
Figure 7.6 Strain profiles of the transverse reinforcement at increasing ductility levels
72
7.2.5 Joint Behavior
As noted in section 6.2, no large cracks or opening appeared during stage 1 of the testing.
However, in examining the data some very small cracks did begin to form at the joints
nearest to the column. The hysteretic behavior of one of these joints is shown in Figure
7.7. The initial state of the joint shown in the figure is after the boundary conditions have
been applied, thus the data does not begin at the graph’s origin. The change in stiffness as
the joint cracked is apparent in the bilinear stiffness of the data. Also a hairline residual
crack on the order of 0.001 in. ( 0.025mm) formed at the joint.
- 0.001 0.000 0.001 0.002 0.003 0.004 0.005 0.006
Opening across joint ( in.)
- 150
- 100
- 50
0
50
100
150
Lateral Force ( k)
Joint 363
0.00 0.05 0.10 0.15
( mm)
- 500
- 100
300
700
( kN)
Figure 7.7 Joint opening during stage 1
73
7.3 Test Stage 2
7.3.1 Force- Displacement Behavior
The stage 2 performance is shown in Figure 7.8, superimposed on the force- displacement
data from stage 1. Similarly to the data from the first stage, the force- displacement
diagram of the system shows a stable hysteresis loop. The flexural capacity of the column
increased due to the higher axial load used in stage 2. Similarly to stage 1, the initial
monotonic prediction was slightly exceeded. After μ= 4 the lateral load increases very little
signifying a nearly perfectly plastic hinge at the column top. This accounts for the little
additional opening of the superstructure joints as further displacement levels were reached.
- 10 - 8 - 6 - 4 - 2 0 2 4 6 8 10
- 250
- 200
- 150
- 100
- 50
0
50
100
150
200
250
- 5 - 4 - 3 - 2 - 1 0 1 2 3 4 5
Drift ratio (%)
- 9 - 8 - 7 - 6 - 5 - 4 - 3 - 2 - 1 0 1 2 3 4 5 6 7 8 9
Stage 2
Stage 2 Prediction
- 1000
- 600
- 200
200
600
1000
Lateral load ( kN)
- 200 - 100 0 ( mm) 100 200
Displacement at top of the column ( in.)
Lateral load ( kips)
Displacement ductility factor
Stage 1
Figure 7.8 Stage 2 force- displacement diagram
74
7.3.2 Curvature
The stage 2 curvature profile is shown in Figure 7.9. The range of curvature greatly
exceeding the yield curvature gradually crept down from the top of the column as higher
ductilities were reached. In the final cycles at μ= 8, the total plastic hinge length appears to
be around 48 in ( 1219 mm).
- 0.0006 - 0.0004 - 0.0002 0.0000 0.0002 0.0004 0.0006 0.0008 0.0010
Curvature ( rads/ in.)
0
24
48
72
96
120
144
Height above footing ( in.)
0
500
1000
1500
2000
2500
3000
3500
Height above footing ( mm)
- 0.02 - 0.01 0.00 0.01 0.02 0.03
Curvature ( rads/ m)
Ductility = 4
Ductility = 5
Ductility = 6
Ductility = 8
Figure 7.9 Stage 2 curvature profile
75
7.3.3 Longitudinal Bar Behavior
The tensile strains along the lengths of the longitudinal bars during stage 2 of the test are
shown in Figure 7.10. For the location of the bars refer to Figure 7.3. Assuming a
theoretical yield strain of 0.005 ( 5000 microstrain) the spread of the plastic hinge down the
column is evident. At μ= 4 the apparent plastic hinge length is 30 in. ( 760mm) as at the end
of testing stage 1, while at μ= 8 it is approximately 48 in ( 1220mm). Although it is difficult
to see because many gages were damaged, the strain penetration into the pier segment also
increased significantly from the 12 in. ( 305mm) at the end of stage 1.
Other gaged bars had few functioning
gages remaining
0 5000 10000 15000 20000
Microstrain ( in/ in or mm/ mm)
80
100
120
140
160
180
Height above footing ( in.)
2000
3000
4000
5000
( mm)
Bar SLS2F
Column Top
0 5000 10000 15000 20000
Microstrain ( in/ in or mm/ mm)
80
100
120
140
160
180
Height above footing ( in.)
2000
3000
4000
5000( mm)
Bar SLS8A
Column Top
0 5000 10000 15000 20000
Microstrain ( in/ in or mm/ mm)
80
100
120
140
160
180
Height above footing ( in.)
2000
3000
4000
5000
( mm)
Bar SLS8F
Column Top
Figure 7.10 Longitudinal bar strain profiles for stage 2 at increasing ductility levels
76
7.3.4 Column- Superstructure Interaction
As seen in the force- displacement data for stage 2 of the test, due to the additional vertical
load applied to the column the capacity and stiffness of the column increased. This
increase in stiffness as well as the additional moment caused by the higher vertical load
caused some opening of the joints nearest the column. Any increase in column stiffness
will cause higher moments to be transferred to the superstructure under similar
displacements. However once the joint opens the flexibility of the entire system increases.
Definitive data from the single test verifying this result was difficult because of the effect
of the increased vertical loads. Further analysis and investigation will be included in the
subsequent report detailing the analysis portion of the project.
One area of the column- superstructure interaction where relevant data was gained was in
the pier segment. During stage 1 of the test there was no significant opening of the joints
nearest to the column and subsequently the gauged bars in the pier segment saw very low
strains. The vertical headed- bars placed at each corner of the pier segment with the intent
of picking up the shear forces ( refer to Figure 2.5) saw strains on the order of 4% of the
yield strain ( εy) with a maximum of 200 microstrain throughout stage 1. During stage 2 of
the test when the joints opened somewhat, the strains in these bars increased to around
25% of εy. The locations of two of these bars as well as their strains during stage 2 are
shown in Figure 7.11.
77
- 6 - 4 - 2 0 2 4 6 8
System Displacement ( in.)
0
200
400
600
800
1000
1200
microstrain
PRK7
- 6 - 4 - 2 0 2 4 6 8
System Displacement ( in.)
0
200
400
600
800
1000
1200
microstrain
PRQ7
- 6 - 4 - 2 0 2 4 6 8
System Displacement ( in.)
0
200
400
600
800
1000
1200
microstrain
PRQ7
Figure 7.11 Strain of corner pier segment headed bars during stage 2
7.3.5 Joint Behavior
Clear opening of the joints nearest the column was seen during stage 2 of the test. The
opening data of the joints on either side of the column is shown in Figure 7.12. The
hysteretic behavior of the joint is shown, with energy dissipation likely do to debonding of
the tendon and spreading of the crack down the sides of the superstructure. The joint loops
are offset likely because of calibration of the instruments or possibly shifting during the
weeks of testing. Figure 7.13 illustrates the differences between stages 1 and 2 at μ= 4. It
can be seen that the stiffness of the joint opening is greatly reduced due to the larger
78
openings in stage 2. Due to the boundary conditions differences between the stages the
graphs appear to cross separate origins.
0.00 0.01 0.02 0.03
Opening across joint ( in.)
- 150
- 100
- 50
0
50
100
150
Lateral Force ( k)
Joint 363
0.0 0.2 0.4 0.6 0.8
( mm)
- 600
- 400
- 200
0
200
400
600
( kN)
Joint 309
Figure 7.12 Joint opening during stage 2
79
0.00 0.01 0.02 0.03
Opening across joint ( in.)
- 150
- 100
- 50
0
50
100
150
Lateral Force ( k)
0.0 0.2 0.4 0.6 0.8
( mm)
- 600
- 400
- 200
0
200
400
600
( kN)
Joint 363
Joint 309
Figure 7.13 Comparison of joint opening at μ= 4 between stages 1 and 2
80
8 Conclusions
8.1 Overview
This report discussed the design and testing of a half- scale bridge system built using the
balanced cantilever method of precast segmental construction. Issues regarding design and
analysis of precast segmental bridges are discussed based on the results from the segmental
bridge system test as well previous phases of the testing program [ 6]. Issues for future
research are highlighted where applicable.
8.2 Joint Performance
Despite not being noticed visibly during stage 1 of the test, the joints began to crack at the
locations of the cast- in- place closure joint. The cracking was only at the top surface of the
superstructure joints. The hairline cracks that formed along the joint were not easy to see,
and closed up completely upon unloading as evidenced by their not being visually
recognized during or immediately after the end of stage 1 testing. This cracking was likely
due to a lack of tensile strength across the joint after decompression of the Prestress force
occurred. At the interface between the pier segment and the CIP closure ( or the first
segment and the CIP closure) it should be assumed that there is no tensile strength from the
concrete at that interface. No other joints were seen to open during any stage of testing.
In the second stage of testing the joints nearest to the column experienced noticeable
opening. The cracks were seen to form at the first peaks of loading, and to open and close
throughout the cyclic loading of the bridge. No residual damage in the joint region
occurred and upon unloading the joints closed again. The crack width did not expand
81
greatly as the system displacements rose from a ductility of 4 to ductility 8, this was due to
the small increase in moment transferred from the column during these cycles as the
column had already formed a plastic hinge. The crack widths ranged from approximately
0.5 to 0.8 mm. The crack was not located completely on one or the other side of the 3 in.
closure gap, but progressed along one or both sides in a non- uniform manner. The opening
and closing of the joints was shown through the data collected to be stable, and to dissipate
some energy as the crack propagated down the sides of the superstructure. The crack was
never seen to lengthen more than halfway down the superstructure sides.
The combined effect of the reduced number of tendons and the increased load caused the
joints to open. An area of future research could isolate these variables to gather specific
experimental data related to the change in stiffness to compliment the analytical work
currently in progress as the continuation of this program.
Areas of interest for future research include the change in behavior of the joints as the
bonded tendons gradually debond. This change in the stiffness of the tendons would have
an effect on the opening of the joints. Also, different design levels that allow more
opening of the joints and spread displacement capacity from the column to the
superstructure through the joints could likely allow for greater displacement capacities
while reducing the residual damage sustained by the bridge system. A more severe
combined longitudinal and transverse loading of the superstructure segments could clarify
possible issues dealing with the shear transfer during extensive opening of the joints.
82
8.3 Column Performance
The overall performance of the column was typical of a well- confined, reinforced concrete
column showing high- energy dissipation capacity, high displacement ductility, and stable
hysteretic response. The response at each ductility level was similar during the first and
second cycles. The amount of damage to the column was limited due to the unidirectional
loading used in the test. Extensive spalling of the cover concrete occurred after μ= 4. The
system was not brought to failure due to stability issue and possible future uses of the test
unit. However failure was certain to occur in the confined corner elements of the column
by rupture of the longitudinal or spiral reinforcement or crushing of the corner element
concrete.
At a displacement ductility level of 4 the plastic hinge length was similar to that predicted.
However, at the extremely high displacement levels ( μ= 8), after significant spalling of the
unconfined concrete had occurred, the plastic hinge length continued to extend down the
column reaching a length approximately 1.5 times longer than the initial prediction. This
was corroborated between both the curvature and longitudinal bar strain data during each
stage of the test.
Further research involving precast segmental columns would compliment the findings of
this bridge system test.
83
8.4 Other Remarks
This test provided two ideas regarding seismic design of precast segmental bridges. First
that it is possible to apply the existing bridge design philosophy of limiting all damage to
the column with this type of bridge provided the prestressing steel and column strengths
are properly designed. Second, when the joints are allowed more flexibility and less
prestressing is used, the behavior of the system is not compromised by minor opening of
the joints. These findings are limited by the fact that the test used only loading in the
longitudinal and vertical directions and that adding transverse loading may cause
additional damage or changes to the system behavior once the joints open. This is a likely
area of future research.
The pier segment design intended to safely transfer shear forces through to the column
once the joints open was accomplished by the vertical headed bars placed at each corner of
the pier segment. These were seen to take significantly more load during the stage of the
test in which the joints open. Other than this modification the pier segment used was
similar to those of CIP bridges. It should be noted that no access opening through the pier
segment was used in the design as this highly complicated the already congested region of
the pier segment. The access holes on the bottoms of the superstructure midspan segments
functioned as the entry point into the center of the box girder sections.
84
9 Appendices
9.1 References
1. R. J. G. MacGregor, M. E. Kreger, and J. E. Breen, Strength and Ductility of a Three-span
Externally Post- tensioned Segmental Box Girder Bridge Model, Prepared for
the Center for Transportation Research, Research Report 1209- 2F, University of
Texas at Austin, April, 1991
2. A. N. A. Hindi, M. E. Kreger, and J. E. Breen, Enhancing the Strength and Ductility
of Post- tensioned Segmental Box Girder Bridges, Prepared for the Center for
Transportation Research, Research Report 1209- 2G, University of Texas at Austin,
April, 1991
3. AASHTO, Standard Specifications for Highway Bridges, 14th Edition, 1989
4. AASHTO, Guide Specifications for Design and Construction of Segmental
Concrete Bridges, 2nd Edition, 1999
5. Caltrans. Seismic Design Criteria. Version 1.1, July 1999.
6. S. H. Megally, M. Garg, F. Seible, and R. K. Dowell, Seismic Performance of
Precast Segmental Bridge Superstructures, Research Report SSRP- 2001/ 24,
University of California at San Diego, CA, May 2002.
7. K. P. Burnell, S. H. Megally, and J. I. Restrepo, Cast- In- Place Hollow Rectangular
Column Test, Preliminary Results Summary. Submitted to Caltrans. May 7th, 2003.
8. ASBI. Recommended Practice for Design and Construction of Segmental Concrete
Bridges. April 2000.
85
9. E. M. Hines, A. Dazio, and F. Seible. Structural testing of the San Francisco-
Oakland Bay Bridge east span skyway piers. Structural Systems Research Project
SSRP 2002/ 01, University of California, San Diego, La Jolla, CA 2002
10. E. M. Hines, F. Seible, R. K. Dowell, and A. Dazio. Structural testing of the San
Francisco- Oakland Bay Bridge east span skyway Structures: Longitudinal Pier
Test. Test Report TR 200- 21, University of California, San Diego, La Jolla, CA
2000
11. M. J. N. Priestley, F. Seible, and G. M. Calvi, Seismic Design and Retrofit of
Bridges. Wiley, New York, 1996.
12. Bond Area Scaling
13. E. M. Hines, F. Seible, and M. J. N. Priestley, Cyclic tests of structural walls with
highly- confined boundary elements, Structural Systems Research Project SSRP
1999/ 15, University of California, San Diego, La Jolla, CA 1999
14. ANATECH Corporation, ANDRIANNA- User’s Guide, Version 1.0, 1999
15. J. B. Mander, M. J. N. Priestley, and R. Park, Seismic Design of Bridge Piers,
Research Report 84- 2, Department of Civil Engineering, University of Canterbury,
Christchurch, New Zealand, 1984
16. M. P. Collins, and D. Mitchell, Prestressed Concrete Structures, Prentice- Hall,
London, 1991
17. M. Menegotto, and P. E. Pinto, " Method of Analysis for Critically Loaded
Reinforced Concrete Plane Frames Including Changes in Geometry and Non-
Elastic Behavior of Elements under Combined Normal Force and Bending," I. A.
B. S. E., Symposium on the resistance and ultimate deformability of structures
acted on by well- defined repeated loads, Lisbon, Portugal, 1973
86
18. E. G. Nawy, Prestressed Concrete: A Fundamental Approach, 2nd Edition, Prentice
Hall, Upper Saddle River, New Jersey, 1996
87
9.2 Construction Photos
Figure 9.1 Reinforcement and placing of the footing and pin
Figure 9.2 Corner spirals with gauges in place
88
Figure 9.3 Placement of corner spirals and bottom of column reinforcement
Figure 9.4 View of wall reinforcement in column
89
Figure 9.5 Column during construction prior to pouring top portion
90
Figure 9.6 Placing forms for pouring top of column
91
Figure 9.7 Pier segment reinforcement
Figure 9.8 Tendon anchor and bottom of superstructure reinforcement cage
92
Figure 9.9 Steel forms used for casting superstructure segments, interchangeable inner form
Figure 9.10 Tendons and reinforcement in segment
93
Figure 9.11 First completed segment ( later scrapped because the bottom blisters are backwards)
Figure 9.12 Unloading and storage of segments awaiting placement
94
Figure 9.13 Placement of 1st pair of segments
Figure 9.14 Continuing placement of 1st pair of segments
95
Figure 9.15 Pouring 3” closure gap
96
Figure 9.16 1st pair of segments in place and epoxied with temporary prestressing bars
97
Figure 9.17 Tendon wedge plate and wedges
Figure 9.18 Stressing the tendon
98
Figure 9.19 Cutting off tendon tails and placement of grout caps
Figure 9.20 2nd pair of segments installed
99
Figure 9.21 After all segments have been installed and top tendons stressed
Figure 9.22 Placement of end blocks and fixtures
100
Figure 9.23 Test during placement of actuators
101
Figure 9.24 Damaged test unit after final testing cycles
Figure 9.25 Spalling at top corners at conclusion of testing
102
Figure 9.26 Column displacement and pin rotation close- up
103
9.3 Additional Data
- 200 - 150 - 100 - 50
Microstrain
- 150
- 100
- 50
0
50
100
150
Load ( kips)
CT93
- 600
- 400
- 200
0
200
400
600
Load ( kN)
- 160 - 110 - 60 - 10 40
Microstrain
- 150
- 100
- 50
0
50
100
150
Load ( kips)
CT255
- 600
- 400
- 200
0
200
400
600
Load ( kN)
- 155 - 130 - 105 - 80 - 55 - 30 - 5 20
Microstrain
- 150
- 100
- 50
0
50
100
150
Load ( kips)
CT309
- 600
- 400
- 200
0
200
400
600
Load ( kN)
- 90 - 70 - 50 - 30 - 10 10 30
Microstrain
- 150
- 100
- 50
0
50
100
150
Load ( kips)
CT417
- 600
- 400
- 200
0
200
400
600
Load ( kN)
- 135 - 110 - 85 - 60 - 35 - 10 15
Microstrain
- 150
- 100
- 50
0
50
100
150
Load ( kips)
CT471
- 600
- 400
- 200
0
200
400
600
Load ( kN)
- 175 - 150 - 125 - 100 - 75 - 50 - 25 0
Microstrain
- 150
- 100
- 50
0
50
100
150
Load ( kips)
CT579
- 600
- 400
- 200
0
200
400
600
Load ( kN)
Figure 9.27 Concrete top surface strains for stage 1
North
CT309
CT255 CT471
Locations of Gauges
CT579
CT93 CT417
104
0 50 100 150 200 250
Microstrain
- 150
- 100
- 50
0
50
100
150
Load ( kips)
CB147
- 600
- 400
- 200
0
200
400
600
Load ( kN)
- 150 - 100 - 50 0 50 100 150
Microstrain
- 150
- 100
- 50
0
50
100
150
Load ( kips)
CB309
- 600
- 400
- 200
0
200
400
600
Load ( kN)
- 100 - 50 0 50 100
Microstrain
- 150
- 100
- 50
0
50
100
150
Load ( kips)
CB363
- 600
- 400
- 200
0
200
400
600
Load ( kN)
- 50 0 50 100 150
Microstrain
- 150
- 100
- 50
0
50
100
150
Load ( kips)
CB471
- 600
- 400
- 200
0
200
400
600
Load ( kN)
Figure 9.28 Concrete bottom surface strains for stage 1
North
CB309
CB363
CB147
CB471
Locations of Gauges
105
- 20 - 15 - 10 - 5 0 5 10 15 20 25
Distance from Column ( ft)
- 0.1
0.1
0.3
0.5
Span Deflection ( in.)
Stage 1
1
- 1
2
- 2
3
- 3
4
- 4
- 6 - 4 - 2 0 2 4 6
( m)
- 20 - 15 - 10 - 5 0 5 10 15 20 25
Distance from Column ( ft)
- 0.5
0.0
0.5
1.0
Span Deflection ( in.)
Stage 2
- 6 - 4 - 2 0( m) 2 4 6
4
- 4
5
- 5
6
- 6
8
- 8
Figure 9.29 Span deflections
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
- 1.5
- 1.0
- 0.5
0.0
0.5
1.0
1.5
Rotation ( rads)
Stage 1 Rotation at base of column
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
- 2
0
2
4
Rotation ( rads)
Stage 2 Rotation at base of column
- 600 - 400 - 200 0 200 400 600 800
( kN)
Figure 9.30 Column bottom rotations
106
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
- 100
100
300
500
700
900
1100
STR5A54
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
- 100
400
900
1400
1900
STR5A69
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
- 100
400
900
1400
1900
2400
STR5A84
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
- 100
400
900
1400
1900
STR5A99
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
- 100
400
900
1400
1900
2400
STR5A113
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
- 5000
0
5000
10000
STR5A128
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
- 100
400
900
1400
1900
2400
STR5A143
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
- 100
400
900
1400
STR5A158
- 600 - 400 - 200 0 200 400 600 800
( kN)
Figure 9.31 Transverse bar strains at location 5A for stage 1
107
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
- 20000
- 15000
- 10000
- 5000
0
STR5A54i
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
700
1200
1700
2200
2700
STR5A69i
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
600
1100
1600
2100
2600
STR5A84i
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
600
1100
1600
2100
2600
STR5A99i
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
0
1000
2000
3000
STR5A143i
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
700
900
1100
1300
1500
1700
STR5A158i
- 600 - 400 - 200 0 200 400 600 800
( kN)
Figure 9.32 Transverse bar strains at location 5A for stage 2
108
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
0
100
200
300
400
STR5F39
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
- 100
100
300
500
STR5F54
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
- 100
400
900
1400
1900
STR5F69
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
- 100
400
900
1400
STR5F84
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
- 100
400
900
1400
1900
2400
STR5F99
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
- 100
400
900
1400
1900
2400
STR5F113
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
- 100
400
900
1400
1900
2400
2900
STR5F143
- 600 - 400 - 200 0 200 400 600 800
( kN)
Figure 9.33 Transverse bar strains at location 5F for stage 1
109
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
0
100
200
300
400
STR5F39i
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
100
300
500
700
900
STR5F54i
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
- 200
300
800
1300
1800
2300
STR5F69i
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
600
1100
1600
2100
STR5F84i
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
600
1100
1600
2100
2600
STR5F99i
- 600 - 400 - 200 0 200 400 600 800
( kN)
- 150 - 120 - 90 - 60 - 30 0 30 60 90 120 150 180
Load ( k)
0
5000
10000
15000
20000
STR5F143i
- 600 - 400 - 200 0 200 400 600 800
( kN)
Figure 9.34 Transverse bar strains at location 5F for stage 2
110
- 0.015 - 0.010 - 0.005 0.000 0.005 0.010 0.015
Shear Displacement ( in.)
- 150
- 120
- 90
- 60
- 30
0
30
60
90
120
150
180
Load ( k)
- 0.4 - 0.2 0.0 ( mm) 0.2 0.4
Stage 1
- 600
- 400
- 200
0
200
400
600
800
( kN)
Figure 9.35 Shear deflection during stage 1
111
- 100 - 50 0 50
TAW255
- 150
- 100
- 50
0
50
100
150
Load ( k)
- 600
- 400
- 200
0
200
400
600
800
Load ( kN)
- 100 - 50 0 50
TBE255
- 150
- 100
- 50
0
50
100
150
Load ( k)
- 600
- 400
- 200
0
200
400
600
800
Load ( kN)
- 100 - 50 0 50
TBE336
- 150
- 100
- 50
0
50
100
150
Load ( k)
- 600
- 400
- 200
0
200
400
600
800
Load ( kN)
- 100 - 50 0 50
TBW309
- 150
- 100
- 50
0
50
100
150
Load ( k)
- 600
- 400
- 200
0
200
400
600
800
Load ( kN)
- 100 - 50 0 50
TBW336
- 150
- 100
- 50
0
50
100
150
Load ( k)
- 600
- 400
- 200
0
200
400
600
800
Load ( kN)
- 100 - 50 0 50
TDW363
- 150
- 100
- 50
0
50
100
150
Load ( k)
- 600
- 400
- 200
0
200
400
600
800 Load ( kN)
- 100 - 50 0 50
TEE147
- 150
- 100
- 50
0
50
100
150
Load ( k)
- 600
- 400
- 200
0
200
400
600
800
Load ( kN)
Figure 9.36 Tendon gauge microstrains for stage 1
North
AW255
BE255
BW309
BE336
BW336
DW363
EE147 EE255
112
- 200 - 100 0 100 200 300 400 500
TBE255
- 150
- 100
- 50
0
50
100
150
Load ( k)
- 600
- 400
- 200
0
200
400
600
800
Load ( kN)
- 200 - 100 0 100 200 300 400 500
TBE336
- 150
- 100
- 50
0
50
100
150
Load ( k)
- 600
- 400
- 200
0
200
400
600
800
Load ( kN)
- 200 - 100 0 100 200 300 400 500
TBW309
- 150
- 100
- 50
0
50
100
150
Load ( k)
- 600
- 400
- 200
0
200
400
600
800
Load ( kN)
- 200 - 100 0 100 200 300 400 500
TBW336
- 150
- 100
- 50
0
50
100
150
Load ( k)
- 600
- 400
- 200
0
200
400
600
800
Load ( kN)
- 200 - 100 0 100 200 300 400 500
TDE417
- 150
- 100
- 50
0
50
100
150
Load ( k)
- 600
- 400
- 200
0
200
400
600
800
Load ( kN)
- 200 - 100 0 100 200 300 400 500
TEE147
- 150
- 100
- 50
0
50
100
150
Load ( k)
- 600
- 400
- 200
0
200
400
600
800
Load ( kN)
Figure 9.37 Tendon gauge microstrains for stage 2
BE255 North
BW309
BE336
BW336
DE417
EE147